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Design and Construction of Driven Pile Foundations

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Foreword

The purpose of this report is to document the issues related to the design and construction of driven pile foundations at the Central Artery/Tunnel project. Construction issues that are presented include pile heave and the heave of an adjacent building during pile driving. Mitigation measures, including the installation of wick drains and the use of preaugering, proved to be ineffective. The results of 15 dynamic and static load tests are also presented and suggest that the piles have more capacity than what they were designed for. The information presented in this report will be of interest to geotechnical engineers working with driven pile foundation systems.

Gary L. Henderson
Director, Office of Infrastructure
Research and Development

Notice

This document is disseminated under the sponsorship of the U.S. Department of Transportation in the interest of information exchange. The U.S. Government assumes no liability for the use of the information contained in this document.

The U.S. Government does not endorse products or manufacturers. Trademarks or manufacturers’ names appear in this report only because they are considered essential to the objective of the document.

Quality Assurance Statement

The Federal Highway Administration (FHWA) provides high-quality information to serve Government, industry, and the public in a manner that promotes public understanding. Standards and policies are used to ensure and maximize the quality, objectivity, utility, and integrity of its information. FHWA periodically reviews quality issues and adjusts its programs and processes to ensure continuous quality improvement.

Technical Report Documentation Page

1. Report No.
FHWA-HRT-05-159
2. Government Accession No. 3. Recipient’s Catalog No.
4. Title and Subtitle
Design and Construction of Driven Pile Foundations– Lessons Learned on the Central Artery/Tunnel Project
5. Report Date
June 2006
6. Performing Organization Code
7. Author(s)
Aaron S. Bradshaw and Christopher D.P. Baxter
8. Performing Organization Report No.
9. Performing Organization Name and Address
University of Rhode Island Narragansett, RI 02882
10. Work Unit No.
11. Contract or Grant No.
DTFH61-03-P-00174
12. Sponsoring Agency Name and Address
Office of Infrastructure Research and Development Federal Highway Administration
6300 Georgetown Pike McLean, VA 22101-2296
13. Type of Report and Period Covered
Final Report January 2003–August 2003
14. Sponsoring Agency Code
15. Supplementary Notes
Contracting Officer’s Technical Representative (COTR): Carl Ealy, HRDS-06
16. Abstract
Five contracts from the Central Artery/Tunnel (CA/T) project in Boston, MA, were reviewed to document issues related to design and construction of driven pile foundations. Given the soft and compressible marine clays in the Boston area, driven pile foundations were selected to support specific structures, including retaining walls, abutments, roadway slabs, transition structures, and ramps. This report presents the results of a study to assess the lessons learned from pile driving on the CA/T. This study focused on an evaluation of static and dynamic load test data and a case study of significant movement of an adjacent building during pile driving. The load test results showed that the piles have more capacity than what they were designed for. At the site of significant movement of an adjacent building, installation of wick drains and preaugering to mitigate additional movement proved to be ineffective. Detailed settlement, inclinometer, and piezometer data are presented.
17. Key Words
Driven piles, heave, CAPWAP, static load test, Boston tunnel
18. Distribution Statement No restrictions.
This document is available to the public through the National Technical Information Service, Springfield, VA 22161.
19. Security Classif. (of this report)
Unclassified
20. Security Classif. (of this page)
Unclassified
21. No. of Pages
58
22. Price

Form DOT F 1700.7 (8-72) Reproduction of completed page authorized

 

Chapter 1. Introduction

Pile foundations are used extensively for the support of buildings, bridges, and other structures to safely transfer structural loads to the ground and to avoid excess settlement or lateral movement. They are very effective in transferring structural loads through weak or compressible soil layers into the more competent soils and rocks below. A “driven pile foundation” is a specific type of pile foundation where structural elements are driven into the ground using a large hammer. They are commonly constructed of timber, precast prestressed concrete (PPC), and steel (H-sections and pipes).

Historically, piles have been used extensively for the support of structures in Boston, MA. This is mostly a result of the need to transfer loads through the loose fill and compressible marine clays that are common in the Boston area. Driven piles, in particular, have been a preferred foundation system because of their relative ease of installation and low cost. They have played an important role in the Central Artery/Tunnel (CA/T) project.

 

ROLE OF DRIVEN PILE FOUNDATIONS ON THE CA/T PROJECT

The CA/T project is recognized as one of the largest and most complex highway projects in the United States. The project involved the replacement of Boston’s deteriorating six-lane, elevated central artery (Interstate (I) 93) with an underground highway; construction of two new bridges over the Charles River (the Leverett Circle Connector Bridge and the Leonard P. Zakim Bunker Hill Bridge); and the extension of I–90 to Boston’s Logan International Airport and Route 1A. The project has been under construction since late 1991 and is scheduled to be completed in 2005.

Driven pile foundations were used on the CA/T for the support of road and tunnel slabs, bridge abutments, egress ramps, retaining walls, and utilities. Because of the large scale of the project, the construction of the CA/T project was actually bid under 73 separate contracts. Five of these contracts were selected for this study, where a large number of piles were installed, and 15 pile load tests were performed. The locations of the individual contracts are shown in figure 1 and summarized in table 1. A description of the five contracts and associated pile-supported structures is also given below.

1. Contract C07D1 is located adjacent to Logan Airport in East Boston and included construction of a part of the I–90 Logan Airport Interchange roadway network. New roadways, an egress ramp, retained fill sections, a viaduct structure, and retaining walls were all constructed as part of the contract. Driven piles were used primarily to support the egress ramp superstructure, abutments, roadway slabs, and retaining walls.

2. Contract C07D2 is located adjacent to Logan Airport in East Boston and included construction of a portion of the I–90 Logan Airport Interchange. Major new structures included highway sections, a viaduct structure, a reinforced concrete open depressed roadway (boat section), and at-grade approach roadways. Driven piles were used to support the boat section, walls and abutments, and portions of the viaduct.

Figure 1. Map. Locations of selected contracts from the Central Artery/Tunnel project. This figure is a map of the area of the Central Artery/Tunnel project with the locations of five contracts indicated. In the center of the map is Boston Inner Harbor. Logan Airport is to the east, or right side, of the map. Downtown Boston is to the west, or left side, of the map. Charlestown is at the upper left of the map. Chinatown is at the lower left of the map. East Boston is at the upper center of the map. Interstate 90 starts at the lower left side of the map, proceeds horizontally across the bottom toward the right side of the map, curves upward through the Ted Williams Tunnel under Boston Inner Harbor, and goes to the west, or left side, of Logan Airport. Interstate 93 starts at the lower left side of the map, proceeds for a short distance to the right and up, then curves back to the left. The locations of five contracts are given. Contracts C07D1 and D2 are located at Logan Airport. Contract C08A1 is located above Logan Airport. Contract C19B1 is located in Charlestown. And contract C09A4 is located in Chinatown, which is southwest of Downtown and at the lower left corner of the map.
Figure 1. Locations of selected contracts from the CA/T project.

Table 1. Summary of selected contracts using driven pile foundations.
Contract Location Description
C07D1 Logan Airport I–90 Logan Airport Interchange
C07D2 Logan Airport I–90 Logan Airport Interchange
C08A1 Logan Airport I–90 and Route 1A Interchange
C09A4 Downtown I–93/I–90 Interchange, I-93 Northbound
C19B1 Charlestown I–93 Viaducts and Ramps North of the Charles River

 

3. Contract C08A1 is located just north of Logan Airport in East Boston and included construction of the I–90 and Route 1A interchange. This contract involved new roadways, retained fill structures, a viaduct, a boat section, and a new subway station.(2) Both vertical and inclined piles were used to support retaining walls and abutments.

4. Contract C09A4 is located just west of the Fort Point Channel in downtown Boston. The contract encompassed construction of the I–90 and I–93 interchange, and the northbound section of I–93. Major new structures included surface roads, boat sections, tunnel sections, viaducts, and a bridge.(2) Piles were used to support five approach structures that provide a transition from on-grade roadways to the viaduct sections. Piles were also used to support utility pipelines.

5. Contract C19B1 is located just north of the Charles River in Charlestown. The contract included the construction of viaduct and ramp structures forming an interchange connecting Route 1, Storrow Drive, and I–93 roadways. Major new structures included roadway transition structures, boat sections, retaining walls, and a stormwater pump station.(2) Piles were used to support the ramp structures that transition from on-grade roadways to the viaduct or boat sections.

 

OBJECTIVESThe overall objective of this report is to document the lessons learned from the installation of driven piles on the CA/T project. This includes review and analysis of pile design criteria and specifications, pile driving equipment and methods, issues encountered during construction, dynamic and static load test data, and cost data for different pile types and site conditions.

SCOPEThis report consists of six chapters, the first of which presents introductory and background information about the contracts where significant pile driving occurred. The second chapter discusses the criteria and specifications used for pile design and construction on the CA/T project. The third chapter documents the equipment and methods used for pile driving. Major construction issues encountered during driving, such as pile and soil heave, are also discussed. The fourth chapter presents the results of pile load tests performed on test piles using static and dynamic test methods, including a discussion of axial capacity, dynamic soil parameters, and pile driving criteria. The fifth chapter presents the unit costs for pile driving and preaugering for the different pile types used, as identified in the original construction bids. Finally, the sixth chapter summarizes the important findings of this study.

 

Chapter 2. Driven Pile Design Criteria And Specifications

This chapter presents the pile design criteria and specifications used on the CA/T project in contracts C07D1, C07D2, C08A1, C09A4, and C19B1. These include information on the types of piles used, capacity requirements, minimum preaugering depths, and testing requirements. The subsurface conditions on which the design criteria were based are also discussed.

 

SUBSURFACE CONDITIONS

Representative soil profiles from each of the contract sites are shown in figures 2 through 6 based on the interpretation of geotechnical borings. (See references 4, 5, 6, 7, 8, 9, and 10)

As shown in figures 2 through 5, the conditions encountered at sites in East Boston (C07D1, C07D2, and C08A1) and in downtown Boston (C09A4) are similar. The subsurface conditions at these locations typically consisted of fill overlying layers of organic silt, inorganic sand or silt, marine clay, glacial soils, and bedrock. The subsurface conditions shown in figure 6 for the C19B1 site in Charlestown, however, were different from the other four sites. Organic soils and marine clays were only encountered to a limited extent at the site. Also, the thickness of the fill layer was greater relative to the other sites.

The physical properties and geological origin of the soils encountered at the contract sites are described below.

Bedrock: The bedrock in the area consists of argillite from the Cambridge formation. The condition of the bedrock varies considerably with location, even within a given site. Evaluation of rock core samples indicates that the rock is typically in a soft and weathered condition and contains a significant amount of fracturing. However, hard and sound bedrock was found at some locations.

Glacial Soils: The glacial soils were deposited during the last glaciation approximately 12,000 years ago. These deposits include glacial till, and glaciomarine, glaciolacustrine, and glaciofluvial soils. Till is characterized by a mass of unsorted debris that contains angular particles composed of a wide variety of grain sizes, ranging from clay-sized particles to large boulders. Glaciomarine or glaciolacustrine deposits generally consist of clay, silt, and sand, whereas glaciofluvial deposits contain coarser grained sand and gravel. The glacial soils are typically dense in nature as indicated by high standard penetration test (SPT) resistance, and the piles were typically terminated in these deposits.

Marine Soils: Marine soils were deposited over the glacial soils during glacial retreat in a quiescent deepwater environment. The marine clay layer, as shown in figures 2 through 5, is the thickest unit in the profile, but was encountered only to a limited extent at the Charlestown site. The clay is generally overconsolidated in the upper portions of the layer and is characterized by relatively higher strengths. The overconsolidation is a result of past desiccation that occurred during a period of low sea level. By comparison, the deeper portions of the clay layer are much softer and penetration of the SPT split spoon can sometimes occur with just the weight of the drilling rods alone.

Inorganic Soils: Inorganic silts and sands are typically encountered overlying the marine soils. These soils were deposited by alluvial processes.

Organic Soils: The organic soils that are encountered below the fill generally consist of organic silt and may contain layers of peat or fine sand. These soils are the result of former tidal marshes that existed along the coastal areas.

Fill Soils: Fill material was placed in the more recent past to raise the grade for urban development. The fill layer is highly variable in its thickness and composition, ranging from silts and clays to sands and gravels. The consistency or density is also variable as indicated by the SPT blow counts. The variability in the fill is attributed to the characteristics of the particular borrow source material and the methods of placement.

 

Figure 2. Graph. Soil profile at the contract C07D1 site as encountered in Boring EB3-5. This figure has three components. The first component, on the left side of the figure, is a graph consisting of data points connected by a line. The x axis is the standard penetration test N value, with N referring to the number of blows to drive a pile. The x axis is a logarithmic scale from 1 to 100. The y axis is the depth in meters and descends from zero to 55. Each data point indicates the number of blows required to drive a pile to that depth from the previous recorded depth, or data point. The second component is a bar chart indicating the depths of various types of soil. The bar chart is beside the graph of data points and connecting line, and uses the same y axis parenthesis depth in meters, descending from zero to 55 end parenthesis. For fill soil, the information provided by the graph and chart is as follows: approximate depth range in meters, zero to 6; approximate number of data points, 4; range of number of blows from preceding data point, 9 to 30. For organic silt, the information provided by the graph and chart is as follows: approximate depth range in meters, zero to 6; approximate number of data points, 1; range of number of blows from preceding data point, 1. For sand, the information provided by the graph and chart is as follows: approximate depth range in meters, 7 to 11; approximate number of data points, 2; range of number of blows from preceding data point, 20 to 30. For marine clay, the information provided by the graph and chart is as follows: approximate depth range in meters, 11 to 44; approximate number of data points, 22; range of number of blows from preceding data point, 1 to 10. For sand parenthesis glaciofluvial end parenthesis, the information provided by the graph and chart is as follows: approximate depth range in meters, 44 to 46; approximate number of data points, 1; range of number of blows from preceding data point, 90. For glacial till, the information provided by the graph and chart is as follows: approximate depth range in meters, 46 to 51; approximate number of data points, 4; range of number of blows from preceding data point, 30 to 100. For bedrock, the information provided by the graph and chart is as follows: approximate depth range in meters, 51 to 55; approximate number of data points, 1; range of number of blows from preceding data point, 100. The third component of the figure indicates the embedment depth of test pile ET2-C2, which is adjacent to Boring EB3-5. The embedment depth is approximately 49 meters.
Figure 2. Soil profile at the contract C07D1 site as encountered in Boring EB3-5.

Figure 3. Graph. Soil profile at the contract C07D2 site as encountered in Boring EB2-149. This figure has three components. The first component, on the left side of the figure, is a graph consisting of data points connected by a line. The x axis is the standard penetration test N value, with N referring to the number of blows to drive a pile. The x axis is a logarithmic scale from 1 to 100. The y axis is the depth in meters and descends from zero to 40. Each data point indicates the number of blows required to drive a pile to that depth from the previous recorded depth, or data point. The second component is a bar chart indicating the depths of various types of soil. The bar chart is beside the graph of data points and connecting line, and uses the same y axis: depth in meters descending from zero to 40. For fill soil, the information provided by the graph and chart is as follows: approximate depth range in meters, zero to 4.5; approximate number of data points, 3; range of number of blows from preceding data point, 2 to 35. For organic silt, the information provided by the graph and chart is as follows: approximate depth range in meters, 4.5 to 6; approximate number of data points, 1; range of number of blows from preceding data point, 5. For sand, the information provided by the graph and chart is as follows: approximate depth range in meters, 6 to 9; approximate number of data points, 2; range of number of blows from preceding data point, 30 to 40. For marine clay, the information provided by the graph and chart is as follows: approximate depth range in meters, 9 to 27; approximate number of data points, 11; range of number of blows from preceding data point, 6 to 11. For silt parenthesis glaciomarine end parenthesis, the information provided by the graph and chart is as follows: approximate depth range in meters, 27 to 38; approximate number of data points, 7; range of number of blows from preceding data point, 40 to 100. For bedrock, the information provided by the graph and chart is as follows: approximate depth range in meters, 38 to 40; approximate number of data points, zero; range of number of blows from preceding data point, not applicable. The third component of the figure indicates the embedment depth of test pile 923, which is adjacent to Boring EB2-149. The embedment depth is approximately 32 meters.
Figure 3. Soil profile at the contract C07D2 site as encountered in Boring EB2-149.

 

Figure 4. Graph. Soil profile at the contract C08A1 site as encountered in Boring EB6-37. This figure has three components. The first component, on the left side of the figure, is a graph consisting of data points connected by a line. The x axis is the standard penetration test N value, with N referring to the number of blows to drive a pile. The x axis is a logarithmic scale from 1 to 100. The y axis is the depth in meters and descends from zero to 70. Each data point indicates the number of blows required to drive a pile to that depth from the previous recorded depth, or data point. The second component is a bar chart indicating the depths of various types of soil. The bar chart is beside the graph of data points and connecting line, and uses the same y axis: depth in meters descending from zero to 70. For fill soil, the information provided by the graph and chart is as follows: approximate depth range in meters, zero to 4; approximate number of data points, 3; range of number of blows from preceding data point, 5 to 30. For silt and sand soil, the information provided by the graph and chart is as follows: approximate depth range in meters, 4 to 10; approximate number of data points, 3; range of number of blows from preceding data point, 10 to 30. For marine clay, the information provided by the graph and chart is as follows: approximate depth range in meters, 10 to 40; approximate number of data points, 20; range of number of blows from preceding data point, 1 to 7. For sand and gravel parenthesis glaciofluvial end parenthesis, the information provided by the graph and chart is as follows: approximate depth range in meters, 40 to 65; approximate number of data points, 15; range of number of blows from preceding data point, 30 to 100. For bedrock, the information provided by the graph and chart is as follows: approximate depth range in meters, 65 to 70; approximate number of data points, 3; range of number of blows from preceding data point, 100. The third component of the figure indicates the embedment depth of test pile ET2-C2, which is adjacent to Boring EB6-37. The embedment depth is approximately 45 meters.
Figure 4. Soil profile at the contract C08A1 site as encountered in Boring EB6-37

Figure 5. Graph. Soil profile at the contract C09A4 site as encountered in Boring IC10-13. This figure has three components. The first component, on the left side of the figure, is a graph consisting of data points connected by a line. The x axis is the standard penetration test N value, with N referring to the number of blows to drive a pile. The x axis is a logarithmic scale from 1 to 100. The y axis is the depth in meters and descends from zero to 25. Each data point indicates the number of blows required to drive a pile to that depth from the previous recorded depth, or data point. The second component is a bar chart indicating the depths of various types of soil. The bar chart is beside the graph of data points and connecting line, and uses the same y axis: depth in meters descending from zero to 25. For fill, the information provided by the graph and chart is as follows: approximate depth range in meters, zero to 3; approximate number of data points, 2; range of number of blows from preceding data point, 5 to 30. For organic silt, the information provided by the graph and chart is as follows: approximate depth range in meters, 3 to 7; approximate number of data points, 2; range of number of blows from preceding data point, 4 to 5. For marine clay, for which the data points are in two unconnected groups, the information provided by the graph and chart is as follows: approximate depth range in meters, 7 to 41; approximate number of data points in first group, 2; approximate number of data points in second group, 11; range in first group of number of blows from preceding data point, 30 to 50; range in second group of number of blows from preceding data point, 1 to 7. For sand and silt parenthesis glaciofluvial end parenthesis, the information provided by the graph and chart is as follows: approximate depth range in meters, 41 to 45; approximate number of data points, 3; range of number of blows from preceding data point, 60 to 100. For bedrock, the information provided by the graph and chart is as follows: approximate depth range in meters, 45 to 50; approximate number of data points, zero; range of number of blows from preceding data point, not applicable. The third component of the figure indicates the embedment depth of test pile 12A1-1, which is adjacent to Boring IC10-13. The embedment depth is approximately 44 meters.
Figure 5. Soil profile at the contract C09A4 site as encountered in Boring IC10-13

Figure 6. Graph. Soil profile at the contract C19B1 site as encountered in Boring AN3-101. This figure has three components. The first component, on the left side of the figure, is a graph consisting of data points connected by a line. The x axis is the standard penetration test N value, with N referring to the number of blows to drive a pile. The x axis is a logarithmic scale from 1 to 100. The y axis is the depth in meters and descends from zero to 25. Each data point indicates the number of blows required to drive a pile to that depth from the previous recorded depth, or data point. The second component is a bar chart indicating the depths of various types of soil. The bar chart is beside the graph of data points and connecting line, and uses the same y axis: depth in meters descending from zero to 25. For granular fill, the information provided by the graph and chart is as follows: approximate depth range in meters, zero to 10; approximate number of data points, 6; range of number of blows from preceding data point, 2 to 50. For sand, the information provided by the graph and chart is as follows: approximate depth range in meters, 10 to 14; approximate number of data points, 3; range of number of blows from preceding data point, 8 to 60. For gravel parenthesis glaciomarine end parenthesis, the information provided by the graph and chart is as follows: approximate depth range in meters, 14 to 18; approximate number of data points, 3; range of number of blows from preceding data point, 90 to 100. For glacial till, the information provided by the graph and chart is as follows: approximate depth range in meters, 18 to 22.5; approximate number of data points, 1; range of number of blows from preceding data point, 100. For bedrock, the information provided by the graph and chart is as follows: approximate depth range in meters, 22.5 to 25; approximate number of data points, 1; range of number of blows from preceding data point, 100. The third component of the figure indicates the embedment depth of test pile IPW, which is adjacent to Boring AN3-101. The embedment depth is approximately 23 meters.

Figure 6. Soil profile at the contract C19B1 site as encountered in Boring AN3-101

DESIGN CRITERIA AND SPECIFICATIONS

The variable fill and compressible clay soils encountered at depth necessitated the use of deep foundations. Driven piles were selected, and design criteria and specifications were developed for their installation, ultimate capacity, and testing. Because the CA/T project was located in Massachusetts, the design criteria were required to satisfy the regulations given in the Massachusetts State building code.(13) The technical content of the State code is based on the 1993 edition of the Building Officials and Code Administrators (BOCA) national building code.

The specifications that were used for each CA/T contract are contained in two documents of the Massachusetts Highway Department (MHD). The first document includes the general requirements for all CA/T contracts and is entitled Supplemental Specifications and CA/T Supplemental Specifications to Construction Details of the Standard Specifications for Highways and Bridges (Division II) for Central Artery (I-93)/Tunnel (I-90) Project in the City of Boston.(14)

The specifications pertaining to individual contracts are covered in a second document concerning special provisions.(15) The special provisions are necessary given the uniqueness of the environmental conditions, soil conditions, and structure types found in each contract. The special provisions present specific details regarding the pile types, pile capacity requirements, and minimum preaugering depths.

Information selected from the specification regarding pile types, preaugering criteria, pile driving criteria, and axial load and test criteria is highlighted below.

Pile Types

Two types of piles were specified on the selected contracts of the CA/T: (1) PPC piles, and (2) concrete-filled steel pipe piles. The PPC piles were fabricated using 34.5- to 41.3-megapascal (MPa) (28-day strength) concrete and were prestressed to 5.2 to 8.3 MPa. The design drawings of typical 30-centimeter (cm)- and 41-cm-diameter square PPC piles are shown in figures 7 and 8, respectively.

To prevent damage to the pile tips during driving in very dense materials, the PPC piles were also fitted with 1.5-meter (m)-long steel H-pile “stingers.” In the 41-cm-diameter PPC piles, an HP14x89 section was used as the stinger. The stingers were welded to a steel plate that was cast into the pile toe, as shown in figure 8. Stingers were used intermittently on the 30-;m-diameter PPC piles, consisting of HP10 by 42 sections.

The concrete-filled steel pipe piles were 31 to 61 cm in diameter, with wall thicknesses ranging from 0.95 to 1.3 cm. The piles were driven closed-ended by welding a steel cone or flat plate onto the pile tip prior to driving. Once the pile was driven to the required depth, the pile was filled with concrete.

A summary of the pile types used on the CA/T is given in table 2, along with the estimated quantities driven. The quantities are based on the contractor’s bid quantities that were obtained directly from Bechtel/Parsons Brinckerhoff. As shown in table 2, the 41-cm-diameter PPC piles were the dominant pile type used, accounting for more than 70 percent of the total length of pile driven.

 

Table 2. Summary of pile types used on the selected CA/T contracts.
Pile Type Estimated Length of Pile Driven (m)
C07D1 C07D2 C08A1 C09A4 C19B1 Total
32-cm pipe
5,550
5,550
41-cm pipe
5,578
5,578
61-cm pipe
296
296
30-cm square PPC
7,969
3,981
792
3,658
2,177
18,577
41-cm square PPC
32,918
19,879
8,406
14,326
6,279
81,808

 

Preaugering Criteria

Preaugering was specified for all piles that were installed in embankments or within the specified limits of adjacent structures. Settlement problems observed at the Hilton hotel (contract C07D1) initiated the use of preaugering to reduce the potential for soil heave caused by pile installation. Soil heave is discussed further in chapter 3. The required depth of preaugering varied depending on the contract and pile location, but ranged from 7.6 to 32.0 m below the ground surface.

Pile Driving Criteria

The specifications required that a Wave Equation Analysis of Piles (WEAP) be used to select the pile driving equipment. The WEAP model estimates hammer performance, driving stresses, and driving resistance for an assumed hammer configuration, pile type, and soil profile. The acceptability of the hammer system was based on the successful demonstration that the pile could be driven to the required capacity or tip elevation without damage to the pile, within a penetration resistance of 3 to 15 blows per 2.5 cm.

The pile driving resistance criteria estimated from the WEAP analysis was also used as the initial driving criteria for the installation of the test piles. Additional WEAP analyses were required for changes in the hammer type, pile type or size, or for significant variations in the soil profile. It was also specified that the WEAP analyses be rerun with modifications to the input parameters to match the results obtained from the dynamic or static load test results. Modifications to the driving criteria could be made as appropriate, based on the results of the pile load tests.

 

Figure 7. Drawing. Typical pile details for a 30-centimeter-diameter precast, prestressed concrete pile. This figure is a complex design drawing containing a variety of details and information, some of it unreadable, about what the drawing’s title and legend call a 12-inch solid square prestressed concrete pile. The date of the drawing is March 16, 1998. The major sections of the drawing are entitled Pile Capacity Calculations, Notes, Elevation View, Section A-A, and Section B-B. The Pile Capacity Calculations section contains difficult-to-read calculations concerning pile capacity. The Notes section contains statements about such matters as concrete strength, wire strand size, reinforcing bars, welding, pile capacity, and splicing. The Elevation View shows a horizontal pile of unstated length wrapped in five sections with number 3 rebar spiral reinforcement. The first and last sections have five turns 2.54 centimeters parenthesis one inch end parenthesis apart. The second and fourth sections have 16 turns 7.62 centimeters parenthesis 3 inches end parenthesis apart. The middle section has an unstated number of turns 22.9 centimeters parenthesis 9 inches end parenthesis apart. Section A-A appears to be a cross section of a trapezoidal-shaped pile with a top dimension of 31.11 centimeters parenthesis 12.25 inches end parenthesis, a bottom dimension of 29.84 centimeters parenthesis 11.75 inches end parenthesis, and side dimensions of 30.48 centimeters parenthesis 12 inches end parenthesis. Section B-B appears to be the same cross section view as Section A-A, but with reinforcing strands and rebar in the interior of the pile. The figure also has sketches of pile pickup, storage, and transportation points.

1 foot = 0.30 m

1 inch = 25.4 mm

Figure 7. Typical pile details for a 30-cm-diameter PPC pile.

 

Figure 8. Drawing. Typical pile details for a 41-centimeter-diameter precast, prestressed concrete pile with stinger. This figure is a complex design drawing containing a variety of details and information, much of it unreadable, about what the drawing’s title and legend call a 40.64 cm parenthesis 16-inch end parenthesis solid square prestressed concrete pile. The date of the drawing is April 1, 2000. The drawing contains six sketches. Two are called Elevation, and each of the two appears to be a view of the unstated length of a horizontal pile wrapped with rebar spiral reinforcement. Each pile is divided into five sections. For the top-most pile, the first and last sections have five turns an unreadable number of inches apart. The second and fourth sections have 23 turns 7.62 centimeters parenthesis 3 inches end parenthesis apart. The middle section has an unstated number of turns 15.4 centimeters parenthesis 6 inches end parenthesis apart. For the other horizontal pile, the first and last sections have five turns an unreadable number of centimeters apart. The second section has 198 turns 7.62 centimeters parenthesis 3 inches end parenthesis apart. The third, or middle, section has an unstated number of turns 15.4 centimeters parenthesis 6 inches end parenthesis apart. The fourth section has 23 turns 7.62 centimeters parenthesis 3 inches end parenthesis apart. The other four sketches, labeled Section A-A, Section B-B, Section C-C, and Section D-D are each a cross section of the same pile. Each shows different details of the interior of the pile. The pile in Sections A-A, B-B, and D-D is trapezoidal in shape, with a top dimension of 41.3 centimeters parenthesis 16.25 inches end parenthesis, a bottom dimension of 40.0 centimeters parenthesis 15.75 inches end parenthesis, and side dimensions of 40.6 centimeters parenthesis 16 inches end parenthesis. The sketch in Section C-C is of an inner rectangular portion of the same pile. The dimensions are 39.4 centimeters parenthesis 15.5 inches end parenthesis by 40.6 centimeters parenthesis 16 inches end parenthesis.

1 foot = 0.30 m

1 inch = 25.4 mm

Figure 8. Typical pile details for a 41-cm-diameter PPC pile with stinger.

 

Axial Load and Pile Load Test Criteria

The required allowable axial capacities that were identified in the special provisions are summarized in table 3. Allowable axial load capacities ranged from 311 to 1,583 kilonewtons (kN). Lateral load criteria were not identified in the selected contracts.

Table 3. Summary of pile types and axial capacity (requirements identified in the selected contracts).
Pile Type Required Allowable Axial Capacity (kN)
32-cm pipe
890
41-cm pipe
1,583
61-cm pipe
311
30-cm square PPC
356–756
41-cm square PPC
534–1,379

 

The axial capacity of the piles was verified using pile load tests, which were specified in section 940.62 of the general specifications.(14) The required ultimate capacities for the load tests were specified by applying a minimum factor of safety of 2.0 to the required allowable values. A factor of safety of 2.25 was specified in contract C19B1, which is consistent with the recommended American Association of State Highway and Transportation Officials (AASHTO) criteria for piles designed and evaluated based only on a subsurface exploration, static analysis, WEAP analysis, and dynamic pile testing.

Dynamic load testing was required for test piles and for a portion of the production piles to monitor driving-induced stresses in the piles, evaluate hammer efficiency and performance, estimate the soil-resistance distribution, and evaluate the pile capacity during initial installation driving and restrikes. A waiting period of 12 to 36 hours (h) was required after pile installation before restrike tests could be performed.

Static load tests were required for test piles to confirm that the minimum specified allowable capacity was achieved and to better estimate or establish higher allowable design capacities. Section 1817.4.1 of the Massachusetts State building code says that the load reaching the top of the bearing stratum under maximum test load for a single pile or pile group must not be less than 100 percent of the allowable design load for end-bearing piles. Therefore, the specifications required that the static load test demonstrate that 100 percent of the design load was transferred to the bearing layer. If any of the test criteria were not met, the contractor was required to perform additional static load test(s).

Chapter 3. Construction Equipment And Methods

This chapter presents a description of the equipment and methods used during pile driving operations at the CA/T project in the selected contracts. This includes a general overview of impact hammers, how a pile is installed, and how to tell when a pile has reached the desired capacity. Construction issues associated with pile driving during this project are also presented. Pile heave was identified as an issue during construction of the arrivals tunnel at Logan Airport, which required a significant number of piles to be redriven. At another site at the airport, soil heave resulting from pile driving caused significant movement of an adjacent building and required changes to the installation process, including preaugering the piles to a depth of 26 m.

 

EQUIPMENT AND METHODS

Impact hammers were used to drive all of the piles for the CA/T project. An impact hammer consists of a heavy ram weight that is raised mechanically or hydraulically to some height (termed “stroke”) and dropped onto the head of the pile. During impact, the kinetic energy of the falling ram is transferred to the pile, causing the pile to penetrate the ground.

Many different pile driving hammers are commercially available, and the major distinction between hammers is how the ram is raised and how it impacts the pile. The size of the hammer is characterized by its maximum potential energy, referred to as the “rated energy.” The rated energy can be expressed as the product of the hammer weight and the maximum stroke. However, the actual energy transferred to the pile is much less a result of energy losses within the driving system and pile. The average transferred energies range from 25 percent for a diesel hammer on a concrete pile to 50 percent for an air hammer on a steel pile.(17)

Three types of hammers were used on the selected contracts: (1) a single-acting diesel, (2) a double-acting diesel, and (3) a single-acting hydraulic. The manufacturers and characteristics of the hammers used in these contracts are summarized in table 4, along with the pile types driven. Schematics of the three types of hammers are shown in figures 9 through 11.

Table 4. Summary of pile driving equipment used on the selected contracts.
Make and Model Type Action Rated Energy (kN-m) Pile Types Driven Contracts Designation
Delmag™
D 46-32
Diesel Double
153.5
41-cm PPC
C07D1
I
HPSI 2000
Hydraulic Single
108.5
41-cm PPC
C07D1, C07D2
II
ICE 1070
Diesel Double
98.5
31-cm PPC, 41-cm PPC, 41-cm pipe
C08A1, C09A4
III
HPSI 1000
Hydraulic Single
67.8
41-cm PPC
C19B1
IV
Delmag D 19-42
Diesel Single
58.0
32-cm pipe
C19B1
V
Delmag D 30-32
Diesel Single
99.9
32-cm pipe
C19B1
VI

A single-acting diesel hammer (figure 9) works by initially raising the hammer with a cable and then releasing the ram. As the ram free-falls within the cylinder, fuel is injected into the combustion chamber beneath the ram and the fuel/air mixture becomes pressurized. Once the ram strikes the anvil at the bottom of the cylinder, the fuel/air mixture ignites, pushing the ram back to the top of the stroke. This process will continue as long as fuel is injected into the combustion chamber and the stroke is sufficient to ignite the fuel.

 

Figure 9. Schematic. Single-acting diesel hammer. This figure contains six sketches that together show how a single acting diesel hammer initiates and maintains pile driving. The first sketch, entitled Tripping, names the parts of the hammer. From top to bottom, the parts are: ram, cylinder, exhaust port, anvil, recoil dampener, striker plate, hammer cushion, and helmet. An additional part, the fuel pump, is identified in the second sketch, entitled Fuel Injection. In the first sketch, Tripping, the ram is in a raised position. In the second sketch, Fuel Injection, the ram is descending. In the third sketch, Compression-Impact, the tip of the ram has reached the anvil. In the fourth sketch, Explosion, an explosion has occurred where the tip of the ram met the anvil, and the ram is ascending. In the fifth sketch, Exhaust, the ram has passed the exhaust port from which exhaust is escaping. And in the sixth sketch, Scavenging, the ram has reached its highest position and is starting to descend.
Figure 9. Single-acting diesel hammer

A double-acting diesel hammer (figure 10) works like the single-acting diesel hammer except that the system is closed at the top of the ram. As the ram rebounds to the top of the stroke, gasses are compressed in the bounce chamber at the top of the hammer. The bounce chamber temporarily stores and redirects energy to the top of the ram, allowing the stroke height to be reduced and the blow rate to be increased. Bounce chamber pressure is monitored during pile driving because it is correlated with hammer energy. The stroke of the hammer, and thus the energy, is controlled using the fuel pump. This is effective for avoiding bouncing of the hammer during the upstroke, which can lead to unstable driving conditions and damage to the hammer.(17)

A single-acting hydraulic hammer (figure 11) uses a hydraulic actuator and pump to retract the ram to the top of the stroke. Once the ram is at the top of the stroke, the ram is released and free-falls under gravity, striking the anvil. An advantage of hydraulic hammers is that the free-fall height, and thus the energy delivered to the pile, can be controlled more accurately.

Figure 10. Schematic. Double-acting diesel hammer. This figure contains five sketches. The first sketch, entitled Tripping, names the parts of the hammer. From top to bottom, the parts are: bounce chamber, bounce chamber port, ram, cylinder, inlet or exhaust scavenge port, anvil, recoil dampener, striker plate, hammer cushion, and helmet. In the first sketch, Tripping, the ram is descending from a raised position. In the second sketch, Compression, the tip of the ram is near the anvil. In the third sketch, Compression-Impact, the tip of the ram has reached the anvil. In the fourth sketch, Exhaust, the ram is ascending and exhaust is venting from the inlet or exhaust scavenge ports. In the fifth sketch, Scavenging, the ram is near the top of its ascent. Figure 11. Schematic. Single-acting hydraulic hammer. This figure contains one sketch in which the parts of a single-acting hydraulic hammer are identified. From top to bottom, the parts are: actuator, adjustable cylinder carriage, cage, lift cylinder, segmented ram weight, shock absorber, anvil, and pile sleeve.

Figure 10. Double-acting diesel hammer

Figure 11. Single-acting hydraulic hammer

In preparation for driving, a pile is first hoisted to an upright position using the crane and is placed into the leads of the pile driver. The leads are braces that help position the piles in place and maintain alignment of the hammer-pile system so that a concentric blow is delivered to the pile for each impact. Once the pile is positioned at the desired location, the hammer is lowered onto the pile butt. A pile cushion consisting of wood, metal, or composite material is placed between the pile and the hammer prior to driving to reduce stresses within the pile during driving.

Once the pile is in position, pile driving is initiated and the number of hammer blows per 0.3 m of penetration is recorded. Toward the end of driving, blows are recorded for every 2.5 cm of penetration. Pile driving is terminated when a set of driving criteria is met. Pile driving criteria are generally based on the following: (1) the minimum required embedment depth, (2) the minimum number of blows required to achieve capacity, and (3) the maximum number of blows to avoid damage to the pile. All information that is associated with pile driving activities (e.g., hammer types, pile types, pile lengths, blow counts, etc.) is recorded on a pile driving log.

A typical pile driving log is shown in figure 12. This particular record is for the installation of a 24-m-long, 41-cm-diameter PPC pile installed at the airport as part of contract C07D2. A hydraulic hammer with an 89-kN ram and a 1.2-m stroke was used. The number of blows per 0.3 m of driving was recorded from an embedment depth of 9.5 m to a final depth of 16.5 m. At a depth of 16.5 m, the hammer blows required to drive the pile 2.5 cm were recorded in the right-hand column of the record. Driving was stopped after a final blow count of 39 blows per 2.5 cm was recorded.

Once a pile has been installed, the hammer may be used to drive the pile again at a later time. Additional driving that is performed after initial installation is referred to as a redrive or restrike. A redrive may be necessary for two reasons: (1) to evaluate the long-term capacity of the pile (i.e., pile setup or pile relaxation), or (2) to reestablish elevations and capacity in piles that have been subject to heave. Both of these issues were significant for the CA/T project, and they are discussed in the next section.

Figure 12. Form. Typical pile driving record. This figure is an example of a completed pile driving record. The title is “Precast, Prestressed Concrete Pile Driving Record.” The top portion of the form contains entries for such identifying information as the contractor, the pile contractor, and the contract number. The top portion also contains sections for hammer data, pile data, and driving data. The middle portion of the form is a table with columns for depth in feet and blows per foot. The depth columns are pre-printed. The first row of the depth column is zero to 0.305 meters parenthesis zero to 1 foot end parenthesis, and the rows proceed in increments of 0.305 meters parenthesis 1 foot end parenthesis to 53.4 meters parenthesis 175 feet end parenthesis. The rows of the blows per foot column are not pre-printed. On this particular form, the blows per foot rows, beginning with the row for which the depth is 9.5 meters parenthesis 31 feet end parenthesis and ending with the row for which the depth is 16.5 meters parenthesis 54 feet end parenthesis, have handwritten entries. On the far right of the table is a section entitled Final Resist, with columns for depth and blows per inch. The depth column is blank and the blows per inch column has handwritten entries. The bottom portion of the form has spaces for remarks and for a signature by a contractor representative.
Figure 12. Typical pile driving record.

CONSTRUCTION-RELATED ISSUES

Pile Heave

Pile heave is a phenomenon where displacement of soil from pile penetration causes vertical or horizontal movement in nearby, previously driven piles. Pile heave generally occurs in insensitive clays that behave as incompressible materials during pile driving.(17) In these soils, the elevation of adjacent piles is often continuously monitored during driving to look for heave. If a pile moves in excess of some predetermined criterion, the pile is redriven to redevelop the required penetration and capacity. From a cost perspective, pile heave is important because redriving piles can require significant additional time and effort.

Pile Layout and Soil Conditions

Of the contracts reviewed, pile heave was an issue during construction of the arrivals tunnel at Logan Airport (contract C07D2). The location of the C07D2 site is shown in figure 1. A plan view of the arrivals tunnel structure showing the pile locations is shown in figure 13. The tunnel structure is approximately 159 m in length and is located where ramp 1A-A splits from the arrivals road. The tunnel was constructed using the cut-and-cover method, and thus a portion of the overburden soil was excavated prior to pile driving.

Figure 13. Drawing. Site plan, piling layout for the arrivals tunnel at Logan Airport. This figure is an irregularly shaped drawing. On the right side is an open area labeled “Arrivals Road.” Proceeding to the left, another blank area branches at a diagonal upwards and is labeled “Ramp 1A.” Surrounding the blank areas are grid-like borders that indicate the locations of approximately 576 piles described in the text. At the bottom of the drawing is the legend “Not to Scale.”
Figure 13. Site plan, piling layout for the arrivals tunnel at Logan Airport

Approximately 576 piles were driven beneath the alignment of the tunnel structure. The piles, consisting of 41-cm-diameter PPC piles, were designed to support a concrete mat foundation in addition to a viaduct located above the tunnel. They were generally installed in a grid-like pattern, with a spacing of approximately 1.2 m by 1.8 m center to center (figure 13).

The general subsurface conditions based on borings advanced in the area prior to excavation consist of approximately 3 to 6.1 m of cohesive and/or granular fill, overlying 1.5 to 3 m of organic silt and sand, overlying 12.2 to 42.7 m of soft marine clay, overlying 0.9 to 2.8 m of glacial silts and sands, underlain by bedrock.(6) Excavation was accomplished into the clay layer, resulting in a clay layer thickness of about 6.1 m at the southeastern end of the structure to around 3.7 m at the northwestern end.

The piles were designed for end bearing in the dense glacial silts and sands, and were preaugered to about the bottom of the marine clay layer to minimize heave and displacement of these soils. The preauger depths were approximately 30 to 70 percent of the final embedment depths of the piles. Preaugering was done using a 46-cm-diameter auger, which is the equivalent circular diameter of the 41-cm square pile. The piles were driven using an HPSI 2000 hydraulic hammer.

Field Observations

Pile heave was monitored during construction by field engineers. As described in the Massachusetts State building code and project specifications, piles identified with vertical displacement exceeding 1.3 cm required redriving. According to field records, 391 of the 576 piles (68 percent) installed required redriving. Of those 391 piles, 337 piles (86 percent) were driven in one redrive event, 53 piles (14 percent) required a second redrive event, and 1 pile required a third redrive event. The impact on the construction schedule or costs was not identified. Despite the use of partial preaugering, a significant portion of the piles showed excessive heave and required substantial redrive efforts. Heave is attributed to the displacement of the underlying glacial soils that were not preaugered.

Pile heave issues were not identified on the other CA/T contracts. Since partial preaugering was used on the majority of these contracts, the difference may be related to the spacing between piles. Table 5 summarizes the pile spacing used on the selected contracts. As shown in table 5, the pile spacing of 1.2 m used at the arrivals tunnel structure is significantly less than the spacing used for structures of comparable size. Therefore, it is anticipated that a pile spacing of greater than about 1.8 m may limit pile heave to within the 1.3-cm criterion.

Table 5. Summary of pile spacing from selected contracts.
Contract Structure Foundation Bent Spacing(m) Pile Spacing (m)
C07D1
Ramp ET
Slab
2.7
2.7
Pile cap
1.4
1.4
Egress Ramps
Pile cap
1.8
1.8
C07D2
Arrivals Tunnel
Pile cap
1.8
1.2
Pile cap
1.8
1.2
Pile cap
1.4
1.2
C08A1
South Abutment
Pile cap
3.05
1.8-2.4
East Abutment
Pile cap
1.1–2.7
1.4–2.6
West Abutment
Pile cap
1.1–2.1
1.4–2.7
C09A4
Utilities
Pile cap
2.0–2.7
1.8
Approach No. 1
Slab
3.7
5.6
Pile cap
1.4
2.6
Pile cap
NA
1.4
Pile cap
NA
1.5
Approach No. 2
Slab
4.57
3.1–4.6
Approach No. 5
Slab
3.7–4.9
2.1–4.3
C19B1
NS-SN
Slab
3.7
4.9
Ramp CT
Slab
3.1
4.6
Ramp LT
Slab
2.9–3.2
2.4–3.1

NA = not applicable or available

 

Soil Heave

Soil heave caused by pile driving was primarily responsible for the significant movement observed at a building adjacent to the construction of the east abutment and east approach to ramp ET at Logan Airport (contract C07D1). Shortly after the start of pile driving, settlement in excess of 2.5 cm was measured at the perimeter of the building and cracking was observed on the structure itself. These observations prompted the installation of additional geotechnical instrumentation, installation of wick drains to dissipate excess pore pressure generated during pile driving, and preaugering of the piles to reduce soil displacement. Despite these efforts, heave continued to a maximum vertical displacement of 8.8 cm. (See references 20, 21, 22, and 23.)

Pile Layout and Soil Conditions

The location of the project in relation to the building is shown in figure 14. The portion of the east approach that is adjacent to the building consists of two major structures, including an abutment and a pile-supported slab. Both structures are supported by 41-cm-diameter PPC piles. The layout of the pile foundation system is also shown in figure 14. The piles for the slab are arranged in a grid-like pattern with a spacing of about 2.7 m center to center. A total of 353 piles support the structures.

 

Figure 14. Drawing. Site plan showing locations of piles, building footprint, and geotechnical instrumentation. At the top of this drawing, extending generally horizontally, is the irregular outline of a structure labeled 'Existing Building.' Just below the building are five deformation monitoring points. The sites of an inclinometer and a multipoint heave gauge are also indicated, as are the sites of three vibrating wire piezometers. Below the building is an irregularly shaped area divided into Phase I and Phase II. Within the area, the locations of 353 piles are indicated. The locations of wick drains above and to the left of the pile area are also indicated.
Figure 14. Site plan showing locations of piles, building footprint, and geotechnical instrumentation.

Prior to construction activities, five deformation monitoring points (DMPs) were installed along the front perimeter of the building closest to the work area. The DMPs consisted of 13-cm-long hex bolts fixed to the building. These points, designated DMP-101 through DMP-105, were monitored for vertical movement. The DMPs were monitored initially by the contractor and subsequently monitored by an independent consultant.

The subsurface conditions based on borings advanced in the area consist of approximately 3 to 4.6 m of fill, overlying 3 to 6.1 m of organic silt and sand, overlying 27.4 to 33.5 m of soft marine clay, overlying 6.1 to 12.2 m of glacial silt and sand, underlain by bedrock. The piles were designed as end bearing piles to be driven into the dense underlying glacial materials. The glacial soils were encountered at depths of approximately 39.6 to 45.7 m below the ground surface and bedrock was encountered at a depth of approximately 48.8 m.

Field Observations (Phase I Pile Driving)

Pile driving for the east approach was executed in two phases. The first phase began on April 5, 1995, and concluded on June 10, 1995. The second phase began on July 13, 1995, and concluded on August 17, 1995. The piles were driven using a Delmag D46-32 single-acting diesel hammer. The extent of the first phase of pile driving is shown in figure 15. This first phase of work was performed no closer than 27.4 m from the building. The majority of the piles for the slab were installed from the west side of the site working toward the east during the periods of April 5 to April 23, and May 15 to June 2. The majority of the piles for the abutment were installed at the west end of the site during the period of April 23 to May 15.

Settlement data obtained by the contractor during the first phase of pile driving are shown in figure 15. On April 21, 1995, after approximately 2 weeks of pile driving on the west side of the site, initial heave displacements of 0.9 and 0.7 cm were measured in DMP-102 and DMP-103, respectively. Notable heave was observed at DMP-101 and DMP-104 on May 1, which registered displacements of 1.3 and 0.8 cm, respectively. An initial heave displacement of 0.4 cm was measured in DMP-105 on May 9. The heave increased steadily to maximum values as pile driving commenced toward the east side of the site.

 

Figure 15. Graph. Settlement data obtained during first phase of pile driving. This figure is a graph. The x axis is the date and ranges from January 25, 1995, to June 28, 1995. The y axis is the vertical heave in centimeters and ranges from minus 2 to plus 6. A double-headed arrow on the graph indicates that Phase I extended from April 5, 1995, to June 10, 1995. Five lines connecting five sets of data points are plotted on the graph, one each for deformation monitoring points 101 through 105. From January 1995 to the beginning of Phase I on April 5, 1995, the plots are, for the most part, horizontal, with a vertical heave between minus 1 and zero centimeters. The five plots slope upward to the right during Phase I. The plot for deformation monitoring point 103 has the highest apogee, a vertical heave of approximately 4.4 centimeters near the end of Phase I in early June 1995. The plot for deformation monitoring point 105 has the lowest apogee, a vertical heave of approximately 1.6 centimeters near the end of Phase I. After the end of Phase I, each of the plots begins sloping downward.
Figure 15. Settlement data obtained during first phase of pile driving.

A summary of the maximum heave values attributed to the first phase of driving is given in table 6. The greatest amount of heave occurred in DMP-103, which was centrally located relative to the pile grid. On June 2, 1995, 1 week before completion of construction, the heave measured in DMPs 101 through 103 began to level off and subside.

Table 6. Maximum building heave (in cm) observed during pile driving.
Construction Phase DMP 101 DMP 102 DMP 103 DMP 104 DMP 105
Phase I
2.5
3.5
4.3
3.8
1.6
Phase II
3.6
4.8
5.3
3.7
1.3

 

As a result of the excessive heave (greater than 2.5 cm) observed in the first phase of pile driving, mitigation measures were implemented for the second phase of work. This was critical considering that the second phase involved driving piles even closer to the building. The geotechnical consultant recommended three approaches to limiting heave based on schedule and cost constraints.(24) These included: (1) installation and monitoring of pore pressures in the clay during driving and adjusting mitigating measures as appropriate; (2) installation of wick drains between the Hilton and the work area to intercept and aid in the reduction of pore pressures beneath the Hilton that may be generated from pile driving; and (3) based on the performance of the wick drains, preauger phase II piles to limit soil displacement.

Field Observations (Phase II Pile Driving)

Prior to the start of the second phase of pile driving, three double-nested vibrating wire piezometers (VWPZ) were installed to measure pore pressures. These piezometers were installed in close proximity to three of the existing deformation monitoring points (DMP-102 through DMP-104). Additional instrumentation was also installed following the start of the second phase of work, including a multipoint heave gauge (MPHG) to measure vertical movement with depth and an inclinometer to measure lateral movement. The locations of the additional geotechnical instrumentation are shown in figure 14.

The second phase of pile driving began on July 13, 1995, and concluded on August 17, 1995. The extent of the work area is also shown in figure 14. Pile driving generally progressed from the west side of the site toward the east. The location of the second phase of work was no closer than 15.2 m from the existing building.

Shortly after the start of driving, 200 wick drains were installed from July 20 to July 28, 1995, around the western and northern perimeters of the work area. The drains were installed through the clay layer at a spacing of 1.2 m center to center.

Settlement data for the second phase of work, shown in figure 16, demonstrate that heave began to increase at DMP-101 through DMP-104 approximately 1 week after the start of pile driving. Based on the review of initial settlement data, preaugering was implemented from August 4, 1995, through the completion of construction. Preaugering was accomplished using a 41-cm-diameter auger to a depth of 26 m, which is approximately 50 to 60 percent of the pile’s final embedment depth. The auger diameter is 11 percent less than the 46-cm equivalent circular diameter for a 41-cm square pile.

As shown in figure 16, heave continued to increase even after preaugering was initiated. Net heave values of 3.3 to 13.5 cm (table 6) were observed from the start of preaugering to the completion of pile driving, resulting in total heave values ranging from 2.6 to 8.8 cm.

Figure 16. Graph. Settlement data obtained during second phase of pile driving. This figure is a continuation of the graph in figure 14. The x axis is the date and ranges from April 5, 1995, to August 23, 1995. The y axis is the vertical heave in centimeters and ranges from zero to 10. One double-headed arrow on the graph indicates that Phase I extended from April 5, 1995, to June 10, 1995. A second double-headed arrow indicates that Phase II extended from July 13, 1995, to August 17, 1995. Five lines connecting five sets of data points are plotted on the graph, one each for deformation monitoring points 101 through 105. The plots begin at approximately May 17, 1995, or in the latter part of Phase I. The first data point for each plot falls within the vertical heave range of 0.8 to 3.3 centimeters. The plots then slope upward until the end of Phase I on June 10, 1995. The plots then slope gradually downward until the beginning of Phase II on July 13, 1995, falling to a vertical heave range of 1.2 to 3.5 centimeters. During Phase II, the plots slope sharply upward, reaching apogees near the end of Phase II on August 17, 1995. The plot for deformation monitoring point 103 has the highest apogee, a vertical heave of approximately 8.6 centimeters. The plot for deformation monitoring point 105 has the lowest apogee, a vertical heave of approximately 2.6 centimeters. A small double-headed arrow indicates that wick drains were installed between approximately July 20 and July 28, 1995. A single-headed arrow indicates that preaugering began on approximately July 17, 1995.
Figure 16. Settlement data obtained during second phase of pile driving.

Data from the multipoint heave gauge showed that the magnitude of the heave was relatively constant within the upper 30 m, as shown in figure 17. However, vertical displacement decreases dramatically below this depth to approximately zero at the bedrock depth of approximately 50 m. The maximum heave of approximately 5.1 cm at a depth of 3 m below the ground surface is also consistent with the maximum value of 5.3 cm recorded at DMP-103.

Figure 17. Graph. Multipoint heave gauge data obtained during second phase of pile driving. This figure consists of a graph of data points and connecting lines, and an adjacent bar chart. The x axis of the graph is vertical heave in centimeters and ranges from minus 1 to plus 7. The y axis is initial depth in meters and descends from zero to 60. Three lines connecting three sets of data points are plotted on the graph, one plot each for data collected on August 8, 11, and 18, 1995. The three plots begin at a vertical heave of approximately zero centimeters at an initial depth of approximately 50 meters. The three plots then rise roughly linearly to an initial depth of 30 meters. At that point, the vertical heave is approximately 1 centimeter for the August 8 plot, 2.5 centimeters for the August 11 plot, and 5.5 centimeters for the August 18 plot. From the initial depth of 30 meters to just below the surface, each plot rises in a roughly vertical fashion. The bar chart adjacent to the graph gives the depths of five layers of soil. The layers and depths are: fill, zero to 6 meters; silt and sand, 6 to 10 meters; marine clay, 10 to 44 meters; glacial soils, 44 to 50 meters; and bedrock, 50 to 60 meters.
Figure 17. Multipoint heave gauge data obtained during second phaseof pile driving.

The excess pore pressures recorded during the second phase of pile driving are presented in figure 17. The six gauges shown in figure 18 correspond to three pairs (55894–55895, 55896–55897, and 55898–55899) located adjacent to DMP-102, DMP-103, and DMP-104, respectively. There was an increase in the excess pore pressures throughout the pile driving, with maximum values ranging from 0.6 to 12.8 m of head with an average of 5.9 m. The greatest head was measured in VWPZ-55896 at a location nearest DMP-103. These data suggest that the wick drains were not effective in dissipating all excess pore pressures generated during pile driving.

 

Figure 18. Graph. Pore pressure data obtained during second phase of pile driving. The x axis of this graph is the date and ranges from June 28, 1995, to August 23, 1995. The y axis is excess pore pressure head in meters and ranges from minus 2 to plus 16. Six lines connecting six sets of data points are plotted on the graph, one each for gauges 55894 through 55899. The plots begin at an excess pore pressure head of zero meters on July 11, 1995. The plots then diverge, but generally increase until the final data points on August 17, 1995. On that date, the final data points range from a low excess pore pressure head of approximately 0.6 meters for gauge 55899 to a high excess pore pressure head of approximately 12.8 meters for gauge 55897.
Figure 18. Pore pressure data obtained during second phase of pile driving.

The inclinometer data that were obtained adjacent to the building are shown in figure 19. These data showed increasing lateral movement in the direction of the building during pile driving. The maximum net lateral deformations were relatively constant with depth within the upper 30 m of the profile. A maximum deformation of approximately 6 cm was recorded at a depth of approximately 34 m. Similar to the vertical deformations, the lateral deformations decreased sharply below this depth to zero at the bedrock depth. These data suggest that the lateral deformations are of the same magnitude and behavior as the vertical deformations.


Figure 19. Graph. Inclinometer data obtained during second phase of pile driving. This figure consists of a graph of data points and connecting lines, and an adjacent bar chart. The x axis of the graph is lateral deformation in centimeters and ranges from zero to 7. The y axis is depth in meters and descends from zero to 60. Three lines are plotted on the graph, one plot each for data collected on August 12, 16, and 18, 1995. The three plots begin at a lateral deformation of approximately zero centimeters at a depth of approximately 50 meters. The three plots then curve to the right and upward, each reaching a maximum lateral deformation at a depth of approximately 34 meters. The maximum deformations are: approximately 3.2 centimeters for the August 12 plot, approximately 4.6 centimeters for the August 16 plot, and approximately 6.0 centimeters for the August 18 plot. From a depth of approximately 34 meters to the surface, each plot curves in an irregular fashion to the left. The lateral deformations at a depth of zero meters are: approximately 2.2 centimeters for the August 12 plot, approximately 2.5 centimeters for the August 16 plot, and approximately 3.2 centimeters for the August 18 plot. The bar chart adjacent to the graph gives the depths of five layers of soil. The layers and depths are: fill, zero to 6 meters; silt and sand, 6 to 10 meters; marine clay, 10 to 44 meters; glacial soils, 44 to 50 meters; and bedrock, 50 to 60 meters.bar chart: fill, silt and sand, marine clay, glacial soils, bedrock

Figure 19. Inclinometer data obtained during second phase of pile driving.

 

SUMMARY

Soil heave was recognized early on as a potential problem and following phase I driving in contract C07D1, several mitigation efforts were initiated. These included installing wick drains to promote rapid dissipation of excess pore pressures and preaugering piles through a portion of the soft clay layer to a depth of 26 m. Additional instrumentation was installed, including piezometers, MPHG, and an inclinometer. Despite these efforts, heave during phase II pile driving continued to increase to a maximum displacement of 8.8 cm. The piezometer data indicate that the wick drains were not effective in rapidly dissipating pore pressures generated during pile driving. The deformation data indicated that soil heave can still occur in piles that are preaugered over a portion of their embedded depth.

Chapter 4. Dynamic And Static Pile Load Test Data

This chapter presents the methodology and results of dynamic and static pile load test data for the selected contracts. At least two static load tests were performed per contract, and the results of 15 tests are presented herein. The Pile Driving Analyzer® (PDA) was also used on these piles for comparison, and analyses were performed periodically during production pile installation. Issues related to design loads and load test criteria are discussed, including factors of safety and load transfer requirements. A comparison is made between the results of the static load tests and the CAse Pile Wave Analysis Program (CAPWAP®) analyses. The CAPWAP data suggest that the quake values generally exceed the values typically recommended in wave equation analyses. A review of the literature is presented to evaluate the significance of this finding. High blow counts recorded during the end of driving also suggest that the majority of the estimated pile capacities from CAPWAP are conservative.

 

LOAD TEST METHODS

Dynamic Load Test Methods

Approximately 160 dynamic pile load tests were performed to evaluate pile capacity, driving stresses, and hammer performance during the installation of test piles and production piles. The data presented in this report were obtained from project files. (See references 25, 26, 27, 28, 29, 30, 31, 32, 33, 34.)

The PDA was used to record, digitize, and processes the force and acceleration signals measured at the pile head. These signals were used to estimate static capacity using the Case Method, a simplified field procedure for estimating pile capacity, as well as the more rigorous CAPWAP. The dynamic load test results discussed in this report are primarily from the CAPWAP analyses. A description of the fundamentals of dynamic testing, including CAPWAP, is presented in Design and Construction of Driven Pile Foundations (Federal Highway Administration (FHWA) report no. FHWA-HI-97-013).(17) The dynamic testing was carried out in general accordance with project specifications section 940.62.C,(14) Dynamic Load Tests, and D4945-89 of the American Society for Testing and Materials (ASTM). D4945-89 is entitled “Standard Method for High Strain Testing of Piles.”(35)

CAPWAP is an iterative curve-fitting technique where the pile response determined in a wave equation model is matched to the measured response of the actual pile for a single hammer blow. The pile model consists of a series of continuous segments and the total resistance of the embedded portion of the pile is represented by a series of springs (static resistance) and dashpots (dynamic resistance). Static resistance is formulated from an idealized elastoplastic soil model, where the quake parameter defines the displacement at which the soil changes from elastic to plastic behavior. The dynamic resistance is formulated using a viscous damping model that is a function of a damping parameter and the velocity.

First, the forces and accelerations acting on the actual pile during initial impact are recorded with a strain gauge and accelerometer mounted at the pile head. The measured acceleration is used as input to the pile model along with reasonable estimates of soil resistance, quake, and damping parameters. The force-time signal at the pile head is calculated using the model and is compared to the measured force-time signal. The soil-resistance distribution, quake, and damping parameters are subsequently modified until agreement is reached between the measured and calculated signals. An example of a comparison between a measured and calculated force signal from one of the test piles is shown in figure 20. Once an acceptable match is achieved, the solution yields an estimate of ultimate static capacity, the distribution of soil resistance along the pile, and the quake and damping parameters.

 

Figure 20. Graph. Example of case pile wave analysis program signal matching, test pile 16A1-1. This figure is a graph comparing the plots of a measured force signal and a calculated force signal. The x axis is time in milliseconds and ranges from zero to 80. The y axis is force in kilonewtons and ranges from zero to 2500. The measured force signal and the calculated force signal are largely overlapping, with peaks at approximately 17 milliseconds parenthesis 2100 kilonewtons end parenthesis, 30 milliseconds parenthesis 700 kilonewtons end parenthesis, 49 milliseconds parenthesis 700 kilonewtons for the calculated force, 600 kilonewtons for the measured force end parenthesis, 63 milliseconds parenthesis 550 kilonewtons end parenthesis, and 68 milliseconds parenthesis 500 kilonewtons end parenthesis. The calculated force signal also has a peak at 42 milliseconds parenthesis 650 kilonewtons end parenthesis.
Figure 20. Example of CAPWAP signal matching, test pile 16A1-1.(33)

 

Static Load Test Methods

Static load tests were performed during the test phase of each contract to verify the design assumptions and load-carrying capacity of the piles. Telltale rods installed at various depths within the piles were used to evaluate the load transfer behavior of the piles with regard to the surrounding soil and bearing stratum. The static tests were carried out in general accordance with project specifications section 940.62.B.4,(14) Short Duration Test, and the ASTM’s D1143-81, which is entitled “Standard Test Method for Piles Under Static Axial Compression Load.”(36) The static load test data presented in this report were obtained from the project files. (See references 37 through 50.)

Static loads were applied and maintained using a hydraulic jack and were measured with a load cell. A typical load test arrangement is shown in figure 21. Reaction to the jack load is provided by a steel frame that is attached to an array of steel H-piles located at least 3 m away from the test pile. Pile head deflections were measured relative to a fixed reference beam using dial gauges. Telltale measurements were made in reference to the pile head or the reference beam using dial gauges. Pile head and telltale deflection data were recorded for each loading increment.

 

Figure 21. Drawing. Typical static load test arrangement showing instrumentation. From top to bottom, the items identified in this drawing are: reaction beam, bearing plate with moment connection, load cell with readout box, mirror with scale parenthesis piano wire as reference line end parenthesis, hydraulic jack with pump, loading plate, instrumentation chair, dial gauges to record telltale movement parenthesis 4 places end parenthesis, dial gauges to measure pile top movement parenthesis 3 places end parenthesis, reference beam, dial gauges mounted with magnetic stands.
Figure 21. Typical static load test arrangement
showing instrumentation.(51)

An excerpt from the loading procedures for short-duration load test section 940.62 is given below(14):

  1. Apply 25 percent of the allowable design load every one-half hour up to the greater of the following: [two alternatives are described; the most general is 200 percent of the design load]. Longer time increments may be used, but each time increment should be the same. At 100 percent of the design load, unload to zero and hold for one-half hour; then reload to 100 percent and continue 25 percent incremental loads. At 150 percent, unload to zero and hold for one-half hour; then reload to 150 percent and continue 25 percent incremental loads. In no case shall the load be changed if the rate of settlement is not decreasing with time.
  2. At the maximum applied load, maintain the load for a minimum of one hour and until the settlement (measured at the lowest point of the pile at which measurements are made) over a one-hour period is not greater than 0.254 mm (0.01 inch).
  3. Remove 25 percent of the load every 15 minutes until zero load is reached. Longer time increments may be used, but each shall be the same.
  4. Measure rebound at zero load for a minimum of one hour.
  5. After 200 percent of the load has been applied and removed, and the test has shown that the pile has additional capacity, i.e., it has not reached ultimate capacity, continue testing as follows. Reload the test pile to the 200 percent design load level in increments of 50 percent of the allowable design load, allowing 20 minutes between increments. Then increase the load in increments of 10 percent until either the pile or the frame reach their allowable structural capacity, or the pile can no longer support the added load. If failure at maximum load does not occur, hold load for one hour. At maximum achieved load, remove the load in four equal decrements, allowing 15 minutes between decrements.

The capacity of the test piles was selected as the greater capacity defined by two failure criteria. The first criteria establishes the allowable design capacity as “50 percent of the applied test load which results in a net settlement of the top of the pile of up to 1.3 cm, after rebound, for a minimum of one hour at zero load.” The second criterion uses Davisson’s criteria as described below.

The Davisson offset limit load criterion was used on the project to define the ultimate capacity, or failure, of the test piles.(52) The ultimate load is interpreted as the point at which the displacement of the pile head meets a limit that is offset to the elastic compression line of the pile. For piles less than 61 cm in diameter, the limit is defined by the following linear relationship:

Equation 1. Movement of pile top. Uppercase S subscript lowercase f equals the sum of uppercase S subscript lowercase e plus 0.38 plus the product of 0.008 times uppercase D. (1)

where,

Sf = Movement of pile top (cm).

D = Pile diameter or width (cm).

Se = Elastic compression of total pile length (cm).

The elastic compression in this case refers to the pile deflection that would occur if 100 percent of the applied load was transferred to the toe of the pile (i.e., zero shaft friction), and is given by the following equation:

Equation 2. Elastic compression of total pile length. Uppercase S subscript lowercase e equals the quotient of the product of uppercase Q times uppercase L divided by the product of uppercase A times uppercase E. (2)

where,

Q = Applied load.

L = Total length of pile.

A = Cross-sectional area of the pile.

E = Modulus of elasticity of the pile.

The average load in the pile at the midpoint between two telltale locations was estimated from the elastic shortening of the pile using the following equation:

Equation 3. Average pile load at midpoint. Uppercase Q subscript lowercase average equals the product of uppercase A times uppercase E times the quotient of the sum of uppercase D subscript 1 minus uppercase D subscript 2 divided by delta uppercase L. (3)

where,

A = Area of pile.

E = Modulus of elasticity of the pile.

D1 = Deflection at upper telltale location.

D2 = Deflection at lower telltale location.

deltaL = Distance between the upper and lower telltale locations.

Both equations 2 and 3 require the modulus of elasticity of the pile. The specifications require that the elastic modulus be determined via compression tests performed on representative concrete samples (ASTM C 469-87a). However, this method is not really applicable to the concrete-filled steel pipe piles. It was common practice on the CA/T project to use the upper telltale and pile head deflections to calculate the modulus of the pile using equation 3. This approach was justified by assuming that any preaugering that was performed prior to pile installation would reduce the shaft friction, especially near the pile head. In some cases, the elastic modulus of the PPC piles was determined from a combination of telltale and compression test data using engineering judgment.

 

LOAD TEST RESULTS

More than 160 dynamic tests were performed on the selected contracts to evaluate pile capacity during both the testing and production phases. Of these 160 tests, the results of 28 tests are presented in this report because they correspond to static load tests on 15 piles. Information about each pile tested is shown in table 7, and pile driving information is presented in table 8.

 

Table 7. Summary of pile and preauger information
Test Pile Name Contract Pile Type Preauger Depth (m) Preauger Diameter (cm)
ET2-C2
C07D1 41-cm PPC
0
NA1
ET4-3B
C07D1 41-cm PPC
0
NA
375
C07D2 41-cm PPC
9.1
45.7
923
C07D2 41-cm PPC
24.1
45.7
I90 EB SA
C08A1 41-cm PPC
NI2
40.6
14
C08A1 41-cm PPC
27.4
40.6
12A1-1
C09A4 31-cm PPC
30.5
45.7
12A2-1
C09A4 31-cm PPC
32.0
45.7
16A1-1
C09A4 41-cm PPC
30.5
45.7
I2
C09A4 41-cm PPC
30.5
40.6
3
C09A4 41-cm pipe
24.4
40.6
7
C09A4 41-cm pipe
24.4
40.6
IPE
C19B1 32-cm pipe
7.6
30.5
IPW
C19B1 32-cm pipe
12.2
30.5
NS-SN
C19B1 41-cm PPC
8.2
40.6

Notes:

1. NA = Not applicable.

2. NI = Data not identified.

 

Table 8. Summary of pile driving information.
Test Pile Name Test Type1 Hammer Type2 Embedment Depth (m) Minimum Transferred Energy (kN-m) Recorded Penetration Resistance (blows/2.5 cm) Permanent Set (cm)
ET2-C2
EOD
I
47.5
NI3
7,7,7
0.36
34DR
58.0
11
0.23
ET4-3B
EOD
II
41.1
NI
8,7,10
0.25
NI
50.8
14
0.18
375
EOD
II
16.8
50.2
12,13,39
0.08
7DR
54.2
> 12
< 0.20
923
EOD
II
32.9
46.1
7,7,7
0.36
7DR
51.5
> 8
0.33
I90 EB SA
EOD
III
46.6
25.8
12,10,10
0.25
1DR
25.8
13
0.20
14
EOD
III
45.4
25.8
10,10,16
0.15
1DR
23.1
21
0.13
12A1-1
EOD
III
41.8
20.7
4,4,5
0.51
1DR
28.6
> 7
> 0.36
12A2-1
EOD
III
38.7
15.3
3,4,4
0.64
1DR
18.6
8
0.33
16A1-1
EOD
III
43.3
24.4
6,7,7
0.36
3DR
17.1
11
0.23
I2
EOD
III
37.2
27.1
4,4,4
0.64
1DR
19.0
5
0.51
3
EOD
III
39.6
57.1
11,12,14
0.18
1DR
49.9
30
0.08
7
EOD
III
38.1
49.8
11,11,11
0.23
3DR
50.2
> 16
< 0.15
IPE
EOD
V
19.5
39.6
5,5,5
0.51
1DR
53.0
7
0.36
IPW
EOD
VI
22.6
43.3
5,5,5
0.51
1DR
59.7
8
0.33
NS-SN
EOD
IV
13.4
27.1
8,15,16
0.15
7DR
24.4
26
0.10

Notes:

1. EOD = End of initial driving, #DR = # daysbefore restrike.

2. Hammer types: I = Delmag D 46-32, II = HPSI 2000, III = ICE 1070, IV = HPSI 1000, V = Delmag D 19-42, VI = Delamag D 30-32.

3. NI = Data not identified.

 

Dynamic Results and Interpretation

Dynamic tests were performed both at the end of initial driving of the pile (EOD) and at the beginning of restrike (BOR), typically 1 to 7 days (1DR, 7DR, etc.) after installation. In most cases, the dynamic tests were performed before the static load tests. Test piles ET2-C2 and ET4-3B, however, were dynamically tested during a restrike after a static load test was performed. The ultimate capacities of the 15 test piles as determined by CAPWAP analysis are summarized in table 9. The table lists when the test was performed, as well as the predicted shaft and toe resistance.

Table 9. Summary of CAPWAP capacity data.
Test Pile Name Test Type1 Recorded Penetration Resistance (blows/2.5 cm) Ultimate Capacity2 (kN)
Shaft Toe Total
ET2-C2
EOD
7,7,7
NI3
NI
NI
34DR
11
(2,028)
(1,219)
(3,247)
ET4-3B
EOD
8,7,10
NI
NI
NI
NI
14
(1,744)
(1,975)
(3,719)
375
EOD
12,13,39
(890)
(3,336)
(4,226)
7DR
> 12
(1,245)
(3,514)
(4,759)
923
EOD
7,7,7
667
1,904
2,571
7DR
> 8
(1,664)
(1,708)
(3,372)
I90 EB SA
EOD
12,10,10
934
712
1,646
1DR
13
(1,156)
(1,112)
(2,268)
14
EOD
10,10,16
(449)
(2,237)
(2,687)
1DR
21
(894)
(1,926)
(2,820)
12A1-1
EOD
4,4,5
685
979
1,664
1DR
> 7
(1,103)
(743)
(1,846)
12A2-1
EOD
3,4,4
316
845
1,161
1DR
8
1,023
431
1,454
16A1-1
EOD
6,7,7
956
1,063
2,015
3DR
11
(983)
(876)
(1,859)
I2
EOD
4,4,4
400
1,130
1,530
1DR
5
1,526
489
2,015
3
EOD
11,12,14
(983)
(2,086)
(3,069)
1DR
30
(1,228)
(1,690)
(2,918)
7
EOD
11,11,11
(80)
(2,740)
(2,820)
3DR
> 16
(983)
(1,984)
(2,962)
IPE
EOD
5,5,5
489
1,334
1,824
1DR
7
645
1,535
2,180
IPW
EOD
5,5,5
778
1,223
2,002
1DR
8
1,290
1,468
2,758
NS-SN
EOD
8,15,16
(583)
(1,806)
(2,389)
7DR
26
(858)
(1,935)
(2,793)

Notes:

1. EOD = End of initial driving, #DR = # days before restrike.

2. Values shown in parentheses denote conservative values.

3. NI = Data not identified.

Many of the capacities are listed in parentheses, which indicates that the values are most likely conservative (i.e., the true ultimate capacity is larger). It is recognized in the literature that dynamic capacities can be underestimated if the hammer energy is insufficient to completely mobilize the soil resistance.(53) Specifically, research has shown that blow counts in excess of 10 blows per 2.5 cm may not cause enough displacement to fully mobilize the soil resistance.(53,54) As shown in table 8, the majority of the piles during restrike exceeded 10 blows per 2.5 cm and are thus likely to be lower than the true ultimate capacity of the piles.

The conservativeness of the CAPWAP capacities in certain piles can be illustrated by comparing the load versus displacement curve at the toe evaluated with CAPWAP to that obtained in a static load test. The toe load-displacement curves from test pile 16A1-1 are shown in figure 22. Blow counts of seven blows per 2.5 cm were recorded for this pile during initial driving. The static load test data shown in figure 22 were extrapolated from the telltale data. As shown in figure 22, the maximum resistance mobilized by the pile toe from CAPWAP is approximately 1060 kN. At least 1670 kN were mobilized in the static load test; however, the ultimate value is actually higher since failure was not reached.

Figure 22. Graph. Load-displacement curves for pile toe, test pile 16A1-1. The x axis is load at pile toe in kilonewtons and ranges from minus 1000 to plus 2000. The y axis is displacement in centimeters and descends from zero to 2.5. The graph has two plots. A dotted line is the plot using the case pile wave analysis program. A solid line with data points is the plot from static load test data. Both plots start at a displacement of zero centimeters and a load at pile toe of zero kilonewtons and slope down to the right, although the plot from the static load test data contains a significant interruption upward and to the left before resuming the descent to the right. The plots show that the maximum resistance by the pile toe from the case pile wave analysis program is approximately 1060 kilonewtons, and the maximum resistance in the static load test is at least 1670 kilonewtons.
Figure 22. Load-displacement curves for pile toe, test pile 16A1-1.

Soil quake and damping parameters obtained from the CAPWAP analyses are summarized in table 10. It is often assumed that the quake values are approximately 0.25 cm in typical wave equation analyses. The toe quake values in this study range from 0.25 to 1.19, with an average of 1.6 cm. Large toe quake values on the order of up to 2.5 cm have been observed in the literature.(55,56) However, the quake values in this study appear to be within typical values.(57)

Table 10. Summary of CAPWAP soil parameters.
Test Pile Name Test Type1 Quake (cm) Damping (s/m)
Shaft Toe Shaft Toe
ET2-C2
EOD
34DR
0.43
0.84
0.72
0.23
ET4-3B
EOD
0.56
0.36
0.89
0.82
375
EOD
0.64
1.19
0.33
0.07
7DR
0.51
0.86
0.23
0.20
923
EOD
0.38
1.14
0.72
0.43
7DR
0.23
0.81
0.46
0.43
I90 EB SA
EOD
0.13
0.89
0.16
0.56
1DR
0.38
0.56
0.69
0.69
14
EOD
0.25
0.76
0.39
0.43
1DR
0.25
0.41
0.59
0.43
12A1-1
EOD
1DR
0.38
0.56
0.75
0.16
12A2-1
EOD
1DR
0.25
0.51
0.49
0.33
16A1-1
EOD
3DR
0.25
0.10
1.41
1.15
I2
EOD
0.25
0.51
0.75
0.26
1DR
0.13
0.25
0.46
0.10
3
EOD
0.48
0.64
0.13
0.10
1DR
0.15
0.56
0.33
0.10
7
EOD
0.23
0.64
0.46
0.10
3DR
0.25
0.36
0.52
0.10
IPE
EOD
0.25
0.69
0.62
0.23
1DR
0.38
0.89
0.59
0.23
IPW
EOD
0.38
0.64
0.43
0.23
1DR
0.25
0.36
0.59
0.20
NS-SN
EOD
0.30
0.91
0.52
0.33
7DR
0.13
0.46
0.72
0.49

Notes:

1. EOD = End of initial driving, #DR = # days before restrike.

2. s/m = seconds/meter.

Comparison of CAPWAP Data

A comparison between the EOD and BOR CAPWAP capacities is shown in figure 23. The line on the figure indicates where the EOD and BOR capacities are equal. Data points that are plotted to the left of the line show an increase in the capacity over time, whereas data that fall below the line show a decrease in capacity. In the four piles (12A2-1, I2, IPE, and IPW) where the soil resistance was believed to be fully mobilized for both the EOD and BOR, the data show an increase of 20 to 38 percent occurring over 1 day. The overall increase in capacity is attributed to an increase in the shaft resistance.

Figure 23. Graph. Case pile wave analysis program capacities at end of initial driving and beginning of restrike. The x axis is the capacity at the end of initial driving in kilonewtons and ranges from zero to 6000. The y axis is capacity at the beginning of restrike in kilonewtons and ranges from zero to 6000. A solid line extends upward and to the right from the origin. The line is at all points equidistant from the two axes; thus it indicates where the end of initial driving and beginning of restrike capacities are equal. Three sets of data points are plotted on the graph. Four points are entitled 'fully mobilized' are generally in the left portion of the graph, and are all above the solid line. Four other points are entitled 'beginning of restrike lower bound' are generally in the center portion of the graph, and three are above the solid line. Five points are entitled 'end of initial driving and beginning of restrike lower bound' are generally in the right portion of the graph, and four are above the solid line.
Figure 23. CAPWAP capacities at end of initial driving (EOD) and beginning of restrike (BOR).

Static Load Test Data

Static load tests were performed on 15 piles approximately 1 to 12 weeks after their installation. The test results are summarized in table 11. In general, two types of load deflection behavior were observed in the static load tests (figures 24 through 27).

Table 11. Summary of static load test data.
Test Pile Name Time After Pile Installation (days) Maximum Applied Load (kN) Maximum Pile Head Displacement (cm)
ET2-C2
13
3,122
1.7
ET4-3B
20
3,558
2.4
375
15
3,447
1.6
923
33
3,447
2.4
I90 EB SA
23
3,781
1.6
14
6
3,105
2.2
12A1-1
30
1,512
1.4
12A2-1
24
1,014
0.5
16A1-1
17
3,612
2.6
I2
6
3,558
1.7
3
9
3,959
2.4
7
10
3,167
2.0
IPE
84
2,384
1.3
IPW
10
2,891
4.1
NS-SN
30
2,535
1.3

Test pile 12A1-1 (figure 24) represents a condition where the axial deflection of the pile is less than the theoretical elastic compression (assuming zero shaft friction). This pile was loaded to 1,557 kN in five steps and at no point during the loading did the deflection exceed the estimated elastic compression of the pile. This behavior is attributed to shaft friction, which reduces the compressive forces in the pile and limits the settlement. The significant contribution of shaft friction is also apparent in the load distribution curve shown in figure 25, which shows the load in the pile decreasing with depth. This behavior is typical of test piles ET2-C2, ET4-3B, I90-EB-SA, 12A1-1, 12A2-1, I2, and 3.

Figure 24. Graph. Deflection of pile head during static load testing of pile 12A1-1. This figure and figure 25 cover a condition in which the axial deflection of the pile is less than the theoretical elastic compression. The condition is attributable to shaft friction, which reduces the compressive forces in the pile and limits settlement. The figure is a graph of a load displacement curve. The x axis is the load in kilonewtons and ranges from zero to 2000. The y axis is the deflection in centimeters and descends from zero to 3.5. The plots of the test data begin at approximately the origin and slope downward to the right. For the most part, the plots of the test data do not exceed, that is, fall below, the pile’s estimated elastic compression, which is plotted as a solid line sloping from the origin downward to the right. The Davisson’s line is parallel to and below the plot of the estimated elastic compression.

Figure 25. Graph. Distribution of load in pile 12A1-1. The figure is a graph of the load distribution from telltales. The x axis is the load in pile in kilonewtons and ranges from zero to 2000. The y axis is the depth below ground surface in meters and descends from zero to 45. Five sets of data are plotted on the graph. Each plot slopes downward to the left, indicating that the load in pile decreases with depth.
Figure 24. Deflection of pile head during static
load testing of pile 12A1-1.
Figure 25. Distribution of load in pile 12A1-1.

Test pile 14 (figure 26) represents a condition where the axial deflection is approximately equal to the theoretical elastic compression. This suggests that more of the applied loads are being distributed to the toe of the pile with less relative contribution of shaft friction. This is apparent in figure 27, which shows negligible changes in the load within the pile with depth. This behavior is typical of test piles 375, 923, 14, 16A1-1, 7, IPE, and IPW.

 

Figure 26. Graph. Deflection of pile head during static load testing of pile 14. This figure and figure 27 cover a condition in which the axial deflection of the pile is approximately equal to the theoretical elastic compression. The condition suggests that more of the applied loads are being distributed to the toe of the pile with less relative contributions by shaft friction. The figure is a graph of a load displacement curve. The x axis is the load in kilonewtons and ranges from zero to 4000. The y axis is deflection in centimeters and descends from zero to 3.5. The plots of the test data begin at approximately the origin and slope downward to the right. A portion of the plots of the test data exceeds, that is, falls below, the pile's estimated elastic compression, which is plotted as a solid line sloping from the origin downward to the right. The Davisson's line is parallel to and below the plot of the estimated elastic compression.

Figure 27. Graph. Distribution of load in pile 14. The figure is a graph of the load distribution from telltales. The x axis is the load in pile in kilonewtons and ranges from zero to 4000. The y axis is the depth below ground surface in meters and descends from zero to 40. Four sets of data are plotted on the graph. Each plot is approximately vertical, indicating that the load in pile changes negligibly with depth.
Figure 26. Deflection of pile head during
static load testing of pile 14.
Figure 27. Distribution of load in pile 14.

Of the 15 static load tests, only one test pile (IPW) was loaded to failure according to Davisson’s criteria. These data are shown in figures 28 and 29. This pile showed a significant increase in the deflection at approximately 2,580 kN, subsequently crossing the Davisson’s line at approximately 2,670 kN at a displacement of around 2.5 cm. The telltale data obtained near the toe of the pile indicated that the pile failed in plunging.

Figure 28. Graph. Deflection of pile head during static load testing of pile IPW. This figure and figure 29 cover a condition in which the test pile was loaded to failure according to Davisson’s criteria. The figure is a graph of a load displacement curve. The x axis is the load in kilonewtons and ranges from zero to 4000. The y axis is deflection in centimeters and descends from zero to 4.5. The plots of the test data begin at approximately the origin and slope downward to the right, joining at a load of approximately 2000 kilonewtons. The joined plot continues a gradual downward slope until a load of approximately 2580 kilonewtons, at which point it turns much more sharply downward, crossing the Davisson' line at a load of approximately 2670 kilonewtons. The Davisson' line begins at a load of zero kilonewtons and a deflection of approximately 0.7 centimeters and slopes downward to the right.

Figure 29. Graph. Distribution of load in pile IPW. The figure is a graph of the load distribution from telltales. The x axis is the load in pile in kilonewtons and ranges from zero to 4000. The y axis is the depth below ground surface in meters and descends from zero to 20. Five sets of data are plotted on the graph. Each plot is approximately vertical from ground surface to approximately 7.5 meters below ground surface, at which point each plot slopes downward to the left.
Figure 28. Deflection of pile head during
static load testing of pile IPW.
Figure 29. Distribution of load in pile IPW.

 

All test piles achieved the required ultimate capacities in the static load tests. The required ultimate capacities were determined by multiplying the allowable design capacity by a factor of safety of at least 2.0, as specified in the project specifications. A slightly higher factor of safety of 2.25 was used in contract C19B1. Three of the 15 static tests did not demonstrate that 100 percent of the design load was transferred to the bearing soils. Two of the piles (12A1-1 and 12A2-1) could not transfer the load to the bearing soils because of the high skin friction (figures 24 and 25). Test pile I2 could not demonstrate load transfer because the bottom telltale was not functioning.

Comparison of Dynamic and Static Load Test Data

The capacities determined by CAPWAP and from the static load tests are summarized in table 12, along with the required ultimate capacities. Of the 15 test piles, only one pile (IPW) was loaded to failure in a static load test. Likewise, only four BOR CAPWAP analyses and eight EOD CAPWAP analyses mobilized the full soil resistance. This means that the true ultimate capacity of the majority of the piles tested was not reached, and this makes a comparison of static load test and CAPWAP results difficult.

Test pile IPW was brought to failure in the static load test. Coincidentally, it is anticipated that the CAPWAP capacities for this pile also represent the fully mobilized soil resistance because of the relatively low blow counts (i.e., < 10) observed during driving. Based on a comparison of all data for pile IPW, its capacity increased by approximately 35 percent soon after installation, yielding a factor of safety of approximately 3.0. Note that this pile was preaugered to a depth of approximately half of the embedment depth. The capacity of 2,669 kN determined in the static load test is slightly less than the restrike capacity of 2,758 kN. However, this difference is partly attributed to modifications that were made to the pile after the dynamic testing, but prior to static testing. These modifications included removal of 0.6 m of overburden at the pile location and filling of the steel pipe pile with concrete, both of which would decrease the capacity of the pile measured in the static load test.

Table 12. Summary of dynamic and static load test data.
Test Pile Name Required Allowable Capacity (kN) Required Minimum Factor of Safety Required Ultimate Capacity (kN) CAPWAP Ultimate Capacity1(kN) Ultimate Capacity From Static Load Test (kN)
EOD BOR
ET2-C2
1,379
2.00
2,758
NI2
(3,247)
(3,122)
ET4-3B
1,379
2.00
2,758
NI
(3,719)
(3,558)
375
1,379
2.00
2,758
(4,226)
(4,759)
(3,447)
923
1,379
2.00
2,758
2,571
(3,372)
(3,447)
I90 EB SA
1,379
2.00
2,758
1,646
(2,268)
(3,781)
14
1,379
2.00
2,758
(2,687)
(2,820)
(3,105)
12A1-1
756
2.00
1,512
1,664
(1,846)
(1,512)
12A2-1
507
2.00
1,014
1,161
1,454
(1,014)
16A1-1
1,245
2.00
2,491
2,015
(1,859)
(3,612)
I2
1,245
2.00
2,491
1,530
2,015
(3,558)
3
1,583
2.00
3,167
(3,069)
(2,918)
(3,959)
7
1,583
2.00
3,167
(2,820)
(2,962)
(3,167)
IPE
890
2.25
2,002
1,824
2,180
(2,384)
IPW
890
2.25
2,002
2,002
2,758
2,669
NS-SN
1,112
2.25
2,504
(2,389)
(2,793)
(2,535)

Notes:

1. Capacities shown in parenthesis denote values that are conservative (dynamic load tests) or where failure was not achieved (static load tests).

2. NI = Data not identified.

Chapter 5. Cost Dataof Driven Piles

This chapter presents a summary of the costs associated with pile driving operations on the CA/T project. The costs presented in this report were obtained directly from the contractor and represent the contractor’s bid estimates identified in the individual contracts. The primary purpose of the cost data is to document the approximate cost of pile driving on the CA/T project; however, the data may also be useful to design engineers for planning purposes.

The contractor’s bid costs for pile driving are summarized in table 13 by pile type. Unless noted, the costs in table 13 do not include costs for preaugering or costs associated with the mobilization or demobilization of the contractor’s equipment. Steel pipe piles had the highest unit costs, ranging from $213 per meter for the 81.3-cm pile to $819 for the 154.9-cm pile. Unit costs for the PPC piles were lower, ranging from $72 to $197 per meter for the 30-cm PPC piles and $95 to $262 per meter for the 41-cm piles. As one would expect, the unit costs tended to decrease with the increasing size of the contract. The contractor’s bid costs for preaugering are summarized in table 14. Preaugering was not performed in contract C07D1, and preaugering costs were not identified in the contract C07D2 bid. As shown in table 14, the additional cost of preaugering ranged from $33 to $49 per meter.

Table 13. Summary of contractor’s bid costs for pile driving.
Contract Pile Type Estimated Length of Pile Installed (m) Estimated Cost of Installation Cost per meter of Pile1
C19B1
32-cm concrete-filled steel pipe
550
$1,183,650
$213.19
C09A4
41-cm concrete-filled steel pipe
5,578
$1,647,000
$295.27 2
C19B1
61-cm concrete-filled steel pipe
296
$242,500
$819.26
C08A1
30-cm square PPC with stinger
792
$156,000
$196.97
C19B1
30-cm square PPC with stinger
2,177
$285,720
$131.24
C09A4
30-cm square PPC
3,658
$600,000
$164.02 2
C07D2
30-cm square PPC with stinger
3,981
$289,510
$72.72
C07D1
30-cm square PPC with stinger
7,955
$652,500
$82.02
C19B1
41-cm square PPC with stinger
6,279
$824,000
$131.23
C08A1
41-cm square PPC with stinger
8,406
$2,206,400
$262.48
C09A4
41-cm square PPC with stinger
14,326
$3,290,000
$229.65 2
C07D2
41-cm square PPC with stinger
19,879
$2,396,800
$120.57
C07D1
41-cm square PPC with stinger
32,918
$3,132,000
$95.15

Notes:

1. Unit costs include the costs of materials and labor for pile driving only. Preaugering is not included unless otherwise noted. See table 14 for preaugering unit costs. Mobilization and/or demobilization costs are not included.

2. Unit costs include the costs of preaugering.

Table 14. Summary of contractor’s bid costs for preaugering.
Contract Preaugering Depth Range (m) Estimated Total Preaugering Depth (m) Estimated Cost of Preaugering Estimated Cost per meter
C08A1
0 to 30.5
2,134
$70,000
$32.80
C19B1
0 to 30.5
3,712
$182,655
$49.21

Chapter 6. Lessons Learned

This chapter presents a summary of the lessons learned from driven piles on the CA/T project. The conclusions presented below are based on the evaluation of field records, project specifications, and pile load test data compiled from the project files. Five contracts were evaluated, including three located in East Boston/Logan Airport, one located in downtown Boston, and one located in Charlestown. Significant findings are summarized below:

  • The dominant pile type used on the CA/T project was a 41-cm square PPC pile. Based on the contractor’s bid estimates, the PPC piles were also the most economical pile type.
  • Pile heave in excess of the 1.3-cm criteria was identified on one cut-and-cover tunnel structure requiring 445 restrike events for the 576 piles used in the structure. The heave occurred even though preaugering of the marine clay layer was performed. Pile heave issues were not identified at other structures where the pile spacing was greater than about 1.8 m.
  • Installation of displacement piles in contract C07D1 caused excessive movement of an adjacent structure. Despite the use of wick drains and partial preaugering, vertical displacement continued up to 8.8 cm. The wick drains were not effective in rapidly dissipating excess pore pressures from pile driving.
  • The heave issues observed in contract C07D1 prompted the use of preaugering on subsequent contracts. Preaugering was performed over a portion, generally 30 to 70 percent, of the final pile embedment depth.
  • Pile capacities evaluated using dynamic methods were conservative in hard driving conditions (i.e., penetration resistance greater than 10 blows per 2.5 cm) where the soil resistance may not be fully mobilized.
  • Quake values from CAPWAP analyses ranged from 0.25 to 1.19 cm, with an average value of 0.64 cm. These values are higher than the values typically used in wave equation analyses; however, they are within the range of published values.
  • Comparison of CAPWAP data evaluated at the end of initial driving and during restrike shows that the capacity of the piles increased over time by at least 20 percent from an increase in shaft resistance.
  • Only 1 out of 15 piles tested in a static load test was brought to failure according to Davisson’s criteria, because the specifications did not specifically require that the pile be brought to failure.
  • Three of the 15 piles did not successfully demonstrate that 100 percent of the design load was transferred to the bearing soils. Two piles did not meet the criteria because of high shaft friction, and the third did not meet the criteria because of a malfunctioning bottom telltale.
  • Comparison of dynamic and static load test capacities was only possible on one pile (IPW), which reached the Davisson’s failure criteria in the static load test. CAPWAP and static capacities were in good agreement for this pile.

(Source: www.fhwa.dot.gov)

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